00133 Massimo Fragiacomo - Timber Design Society

SEISMIC ANALYSIS OF A LIGHT-FRAME TIMBER BUILDING
WITH AND WITHOUT FRICTION PENDULUM BASE
ISOLATION
Massimo Fragiacomo 1 , Claudio Amadio 2 , Ljuba Sancin 3 , Giovanni Rinaldin 4
ABSTRACT: Light-frame construction is used extensively in the European, American and Australasian market for low
and medium rise timber buildings. These buildings are light-weight and have a high dissipative capacity, and thus a
high behaviour factor can be used in seismic design. On the other hand the use of a high behaviour factor may imply
significant structural and non-structural damage at the end of a high intensity earthquake, leading to potentially
significant economic losses. Passive base isolation is by far the most effective way to reduce the effect of an earthquake
on a structure. In this paper, the use of passive base isolation is investigated for light-frame multi-storey timber
buildings. Two designs of a three-storey light-frame building were carried out according to Eurocode 5 and the Italian
regulation for construction: one with and the other without Friction Pendulum System isolators. The seismic
performance and cost of both solutions are compared, demonstrating the convenience of using passive base isolation.
KEYWORDS: Light-frame structure, cyclic modelling, seismic isolation, timber building
1 INTRODUCTION 123
According to the modern philosophy of the Damage
Avoidance Design [1], a structure should be designed
not only to survive a high intensity earthquake but also
to minimize the structural and non-structural damage.
Notoriously, passive base isolation is by far the most
effective way to reduce the consequences of an
earthquake on an existing or new structure [2] [3] [4]. Its
use has been somewhat limited by the cost, which is
often believed to be high. However, significant progress
has been made in this decade on passive base isolation,
and nowadays the cost does not seem to be an issue
anymore. For these reasons, passive base isolation
system was recently used also for residential
applications, for example in L’Aquila (Italy) where both
1 Massimo Fragiacomo, Associate Professor, Department of
Architecture, Design and Urban Planning, University of
Sassari, Palazzo del Pou Salit, Piazza Duomo 6, 07041-
Alghero, Italy. E-mail: fragiacomo@uniss.it
2 Claudio Amadio, Associate Professor, Department of Civil
Engineering and Architecture, University of Trieste, Piazzale
Europa, 1, 34127 Trieste, Italy. E-mail: amadio@units.it
3 Ljuba Sancin, Former graduate student, Department of Civil
Engineering and Architecture University of Trieste, Piazzale
Europa, 1, 34127 Trieste, Italy . E-mail:
ljuba.sancin@gmail.com
4 Giovanni Rinaldin, PhD student, Department of Civil
Engineering and Architecture University of Trieste, Piazzale
Europa, 1, 34127 Trieste, Italy . E-mail:
giovanni.rinaldin@phd.units.it
reinforced concrete and timber buildings were isolated.
Simple design procedures have been proposed for
passive base isolation systems [5], and extensive design
provisions are available in codes of practice such as the
Eurocode 8 [6]. The effectiveness of passive base
isolation was demonstrated by the earthquakes occurred
in Los Angeles in 1994 and in Kobe in 1995, where a
dramatically different performance was observed in
isolated compared to non-isolated buildings. It is
therefore of interest to investigate this option also for
timber buildings, where the significantly reduced weight
compared to reinforced concrete allows the use of
smaller and, therefore, potentially cheaper base devices.
The paper explores the use of passive base isolation with
Friction Pendulum System (FPS) devices for light-frame
multi-storey timber buildings. Two designs of a threestorey
light-frame building were carried out according to
Eurocode 5 [7] and the Italian regulation for construction
[8]: one with and the other without isolators. The seismic
performance and cost of both solutions are compared,
demonstrating the advantage of passive base isolation.
2 DUCTILITY IN LIGHT-FRAME
TIMBER STRUCTURES
Light-frame timber structures are composed by plywood
or other types of sheathing nailed on light timber frames.
It is well-known that the dissipative behaviour of a lightframe
timber structure is mainly governed by the panelto-frame
connection in the first loading cycles. The

collapse can occur for localized splitting of the wood
(brittle) or for connection failure (ductile).
2.1 OVERSTRENGTH OF THE CONNECTION
In the design of a light-frame building, the wood
elements cannot be considered as ductile. For this
reason, Capacity Based Design is applied to make sure
that the expected ductility in the connection will be
attained before brittle failures of the wood members
occur [9]. The strength capacity of the ductile element
must be smaller than the strength capacity of the brittle
element. In this regard it is particularly important to
evaluate the overstrength of the connection for the
correct design of the connected wood elements. It should
be pointed out that the characteristic and not the design
value of the strength (namely, a material partial factor
γ
M
1) should be used in ductile design of connections,
as prescribed by the Eurocode 8 [6] and by the New
Zealand Timber Standard [10]. The use of the design
values would lead to overdesign the connections and that
could postpone if not prevent at all the plasticization
under the design seismic actions, in this way achieving
no energy dissipation.
2.2 CAPACITY BASED DESIGN CRITERIA
According to a correct design approach in light-frame
timber buildings the only dissipative elements are the
panel-to-frame nailed connections of the shear walls [11]
[12]. All other connections (hold-downs, tie-downs,
angular brackets), all timber members (studs, plates,
sheathing, joists), and the panel-to-joist connections of
the floor diaphragms have to remain elastic and be
designed for the overstrength of the dissipative elements.
Concerning the choice of the ductile collapse
mechanism, little information is provided in the current
Italian [8] and European [6] regulations. A literature
reference [12] suggests that plasticization should be
allowed only at the first floor or at the first and second
floor for buildings up to 4 or from 4 to 6 storeys high,
respectively, whereas all other floors should be designed
for the overstrength of the first floors. In this paper a
similar approach was followed in the design of the 3-
storey case study building. All timber elements and all
nailed panel-to-joist connections were designed for the
overstrength of the panel-to-frame connections of the
shear walls at the first floor. In particular, the nailed
panel-to-frame connections of the shear walls were
calculated at the first floor, where the seismic strength
demand is larger, and reproduced at the upper floors,
where the seismic forces decrease, in order to ensure
plasticization only occurs at the first floor. The brittle
elements including the panel-to-joist floor connections
were therefore designed to resist the forces described in
Equation (1):
R
d, wood
Sd,
wood
Rd
(1)
where
R ,
is the design strength capacity of the timber
d wood
element;
S
d , wood
is the seismic strength demand of the timber
element;
Rd
is the overstrength factor:
Rd
1. 3 according to the
Italian regulation;
S is the overdesign factor of
R
d , connection d , connection
the dissipative connection (at the first floor).
3 CASE STUDY BUILDING
3.1 STRUCTURAL DESCRIPTION
The analysed structure is a three-story light-frame timber
building with two symmetric modules, as can be seen in
Figure 1, connected to each other just at the foundation
and at the top level.
Figure 1: Plan view of the building. The results of the
analysis for the circled wall are displayed in Figure 12
and Figure 13 (dimensions in cm)
The building is regular and is designed as strategic being
assumed as headquarter of the fire brigade. It is situated
in L’Aquila (Italy), 700 m above sea level, in a suburban
zone. Figure 2 displays a cross-section of the building.
Figure 2: Cross-section of the 3-storey light-frame
building (dimensions in cm)
The vertical load-bearing structure is made from lightframe
shear walls with double OSB sheathing (both
sides) nailed on the timber frame. The studs are spaced
60 cm or less and the walls are reinforced with additional
studs where concentrated forces are applied. In the larger

openings, the top plate is integrated with a lintel. The
shear walls are connected to the reinforced concrete
ribbed slab with hold-downs and angle brackets, and to
the upper wall with tie-downs and angle brackets. The
structural flooring is made of two 15 mm thick OSB
sheathings nailed on top and bottom of joists and
blockings. The roof is double pitched, ventilated and
covered with tiles. Walls and floor diaphragms ensure a
box behaviour that can easily carry wind load and
earthquake action. The floor diaphragms and the roof are
regarded as in-plane rigid as recommended by codes of
practice [6] [10] and literature references [12] for timber
floors with sheathings nailed on joists and solid
blockings without significant opening. The light-frame
shear walls provide the lateral load resisting system –
each wall resists the lateral load in the direction parallel
to its plane.
3.2 DESIGN DATA
The main design data of the building are, according to
the Italian regulation [8]:
Permanent load on floors: G k =1.85 kN/m 2
Permanent load on roof: G k =1.60 kN/m 2
Imposed load on floors: Q k =3 kN/m 2 (offices)
Snow load at the ground: q s = 1.27 kN/m 2
Total drag load on windward and leeward
facades: q w =1.18 kN/m 2
Reference service life V r : 100 years (which
affects the design peak ground acceleration)
Site coordinates: 13.3944 E; 43.3660 N
Ground type: B (Deposits of very dense sand,
gravel or very stiff clay, at least several tens of
metres in thickness, characterised by a gradual
increase of mechanical properties with depth)
Peak ground acceleration for CLS (Collapse
Limit State) seismic design: a g = 0.418 g, g
signifying the gravity acceleration
Peak ground acceleration for LLS (Life-safety
Limit State) seismic design: a g =0.331 g
Peak ground acceleration for DLS (Damage
Limit State) seismic design: a g =0.142 g
Peak ground acceleration for OLS (Operational
Limit State) seismic design: a g =0.113 g
Behaviour factor q: q = 4 (Building A – non
isolated); q = 1.5 (Building B – isolated)
The load combination F for ULS and SLS design of the
building subjected to the seismic actions is written in
Equation (2):
F
2
E G k
Q k
(2)
where
2
0. 3 for offices and E is the seismic action,
calculated as a combination of the effects of the
earthquake in X and Y directions (±1.00∙ E x ±0.30∙ E y or
±1.00∙ E y ±0.30∙ E x ). E x and E y vary depending on the
limit states, and are calculated proportionally to the
seismic mass M, which is defined in the Italian
regulation [8], as in Equation (3):
M ( G
2
Q g
(3)
k k
)
The accidental eccentricity is taken into account by
multiplying the seismic effect by the factor
1
0.6 x / L e
where x denotes the distance of the shear wall from the
geometric centre of the floor, in the direction
perpendicular to the earthquake direction, and L e
signifies the distance between the two most spaced shear
walls, measured in the same way as x.
According to the Italian regulation [8], strategic
buildings have to satisfy the requirements of four
different limit states: Collapse (CLS), Life safety (LLS),
Damage (DLS) and Operational Limit State (OLS).
Every limit state has its own return period of the seismic
event and thus the value of the design earthquake
decreases going from the CLS to the OLS. In Table 1 the
values of the base shear forces due to wind and
earthquake in the different building directions are
compared.
Table 1: Comparison among base shear forces due to
wind and earthquake
Building
A
Building
B
Wind X
[kN]
Seismic action X
[kN]
ULS
(LLS)
281.8 740.5
DLS 827.3
ULS
(LLS)
281.8 331.4
DLS 96.8
4 SEISMIC ISOLATION
After designing the non-isolated building, the same
building was re-designed with passive base isolation
(Building B). The base floor diaphragm connecting the
two modules of the building and supported by the
isolation devices is made of a ribbed reinforced concrete
slab. The isolation devices are supported by stiff
concrete columns connected to a flat plate foundation. In
this way it is possible to obtain a basement that can be
used as a garage, as typically required by architectural
considerations. At the same time, according to the Italian
regulation [8] the compulsory requirement of ensuring
an easy maintenance and substitution of the isolation
devices can be satisfied.
4.1 FRICTION PENDULUM SYSTEM DEVICES
Friction Pendulum System devices (FPS) have been used
in this paper as base isolators [4]. They are curved
surface sliding isolators which exploit gravity for recentring,
like a pendulum. Dissipation of the input
seismic energy takes place due to the friction on the
main surface. The parameters of their cyclic behaviour
depend on the curvature and on the friction coefficient.
After the activation, the device develops a lateral force
F din that is described in Equation (4) and is given by the
resultant of the dynamic friction force and the restoring
force due to gravity:

F
din
W
u dinW
R
(4)
where:
F din is the lateral force developed by the isolation device
in action;
W is the weight of the timber superstructure and base
floor diaphragm on the device;
is the dynamic friction coefficient;
din
R is the radius of curvature of the isolator concave
surface;
u is the device displacement.
The curvature radius (see Figure 3) is an important
parameter in these devices, as it governs the stiffness and
therefore the natural vibration period of the isolated
structure. The period T of a rigid mass supported by an
isolator FPS can be calculated as in Equation (5) [13]:
T 2
1
1
g(
R u
din
d
)
(5)
where g is the gravity acceleration and u d is the design
displacement.
The equivalent lateral stiffness k e of an active FPS
device, relative to the maximum displacement u max , is
inversely proportional to the curvature radius R, as can
be seen in Equation (6) [13]:
k e
1
W
R umax
(6)
where the symbols have the same meaning as before.
The direct proportion of the device stiffness to the
weight bearing on it is fundamental for avoiding
torsional problems in the seismic response of the
structure. In this way the stiffness centre coincides
automatically with the mass centre.
4.2 DESIGN OF THE ISOLATION DEVICES
The isolated building (Building B) is designed for the
same gravity and wind load as Building A, but the
seismic forces are smaller. Indeed, the isolation system
was designed to obtain a natural vibration period
T 1 =3.46 s, corresponding to the final part of the design
response spectrum, where the accelerations are less
burdensome. As a result of that, the superstructure of
Building B is lighter, with smaller cross-sections and
with less nail connectors compared to Building A.
The design of the isolators is carried out so as to comply
with two performance requirements: (i) the isolator
should work (move) for at least the CLS and LLS design
levels of earthquake ground motion; and (ii) the isolator
should behave like a fixed restraint for the design wind
load at ultimate limit state. The first condition ensures
that the isolators prevent any structural damage,
according to the Damage Avoidance Design philosophy.
The second design condition is fundamental for the
living comfort, as the wind occurs quite often and the
waving motion of the isolators may cause seasickness of
the occupants.
Figure 3: Cross-section of the FPS device, with double
curved surface [13]
Moreover, the continuous functioning of the isolators
would lead to a major usury of the devices and reduced
safety level in the case of an earthquake.
The displacement demand of the isolation system was
calculated to be over 300 mm at CLS, so that the chosen
device was a double curved surface FPS, with the
following properties:
Vertical load-carrying capacity: 1000/1800 kN
Radius of the sliding surface: 3715 mm
Dynamic friction coefficient: 0.025
Maximum displacement: 350 mm
Equivalent viscous damping coefficient
(corresponding to the maximum design
displacement): 13.4%
Natural period (corresponding to the maximum
design displacement): 3.44 s
Diameter (without anchorage): 620 mm
Depth (without anchorage): 112 mm
The FPS device is displayed in Figure 3.
Figure 4: Positioning and labelling of the isolators
The positioning of the FPSs is shown in Figure 4. The
maximum distance between the devices is almost 8 m.
The limiting factor in deciding the maximum distance
was the bending resistance of the concrete slab ribs and
not the capacity of the isolation devices, which is much
larger than necessary. The governing design conditions
for the concrete slab ribs were sagging bending caused
by gravity loads and also hogging bending at the
supports due to the need to lift the beams about 1.5 cm
for replacing the isolators at the end of their service life.
5 NUMERICAL ANALYSIS
Some non-linear time-history analyses were performed
with the aim to carry out a more accurate comparison of
the seismic performance of both isolated and nonisolated
buildings. The SAP2000 non-linear finite
element program [14], which is a software package

widely used within the design community, was
employed. Since the seismic performance of the whole
building is markedly affected by the behaviour of the
shear walls, careful consideration was given to the
choice of a proper model which can effectively account
for the most important features of the shear wall
behaviour.
5.1 FE MODELLING OF LIGHT-FRAME SHEAR
WALLS
The behaviour of light-frame timber shear walls is fairly
complex as they are composed by different elements:
timber frame, sheathing and connections (panel-to-frame
nailed connection and anchoring of the wall to the
foundation or to the floor underneath). All these
elements contribute to the total flexibility of the wall,
which affects the distribution of horizontal (seismic and
wind) forces within the building. The need to model an
entire 3-storey building has suggested the opportunity to
use a macro-model where all flexibility components are
lumped together, rather than developing a
computationally demanding schematization where all
flexible elements (for example, each nail of the frame-topanel
connection) are explicitly modelled. The macromodel
is displayed in Figure 5: it is made of uniaxial
elements available in the SAP2000 Library. More
specifically, the two lateral studs (chords) and the top
and bottom plates are modelled with rigid beam elements
pinned to each other. The axial flexibility of the chords,
the shear flexibility of the sheathing, and the shear
flexibility of the nailed panel-to-frame connection are
modelled using two diagonal springs, with linear or
cyclic behaviour (NLLink elements in SAP2000),
depending on the type of analysis (linear or non-linear)
carried out. This macro-model has the advantage of
simplicity and low computational demand, and is
suitable for analyses of entire buildings. However, since
the diagonal springs are fictitious elements, their
mechanical properties need to be calibrated on
experimental and analytical results to ensure the
behaviour of the macro-model is equivalent to that of a
real shear wall.
The elastic axial stiffness K diag , the yielding force F y,diag
and the ultimate displacement u ult,diag of each diagonal
can be calculated from simple geometrical
considerations based on the elastic lateral stiffness
K=V/∆ w , the shear force at yielding V y , and the ultimate
horizontal top displacement ∆ w,ult of the wall,
respectively, with the Equations (7), (8) and (9):
K
1 F
2 u
1 V
2
cos( )
1 1
K
2
cos( )
2 cos (
diag
diag (7)
w, diag
w
)
F
y,
diag
1 Vy
(8)
2 cos( )
u cos(
)
(9)
ult, diag w,
ult
where:
F diag is the axial force in one diagonal
u w,diag is the shortening/elongation of the diagonal due to
F diag
V is the lateral force (total shear force) applied on top of
the panel
∆ w is the horizontal top displacement of the wall under
the force V
α is the angle between the plate and the diagonal, as
shown in Figure 5.
The in-plane interstorey deflection of a light-frame shear
wall ∆ w loaded on top can be calculated using the
formulas prescribed by the New Zealand Standard [10]:
∆ w =∆ 4 +∆ 5 +∆ 6 +∆ 7
The four components of flexibility are displayed in
Figure 6.
Figure 6: Deformation components of light-frame timber
shear walls
Figure 5: Deformed configuration of the macro-model of
the shear walls under lateral force applied on the top
5.2 VALIDATION OF THE FE MODELLING
The simplified macro-model was compared with the
analytical formulas and with a more refined FE model
implemented in SAP2000. In this refined model, every
nail was schematized with an elastic spring located in the
actual position with stiffness K ser calculated according to
the EC5 [7] formulas. The OSB panel was modelled with
shell elements and the frame with beam elements. The
top wall displacement due to a 40 kN force applied on
the top of the wall was found to be 0.0073 m using the
refined FE model and 0.0081 m using the macro-model
calibrated on the analytical formulas of Δ 6 , which is the
most important deflection component (the nail slip one).
This comparison demonstrates the accuracy of the
proposed macro-model.

Building A (without isolation) the nodes of the walls
were pinned-connected to the base. In Building B (with
isolation) the cross sections were different and so were
the stiffnesses of the walls.
Figure 7: Calibration of the parameters of the Pivot Cycle
on the experimental results of tests carried out at the
University of Trieste in 2005
Since the FE macro-model was used to model the cyclic
behaviour of a shear wall in time-history analyses, the
“Multilinear plastic - Pivot Cycle” available in the
SAP2000 library was chosen to schematize the hysteretic
behaviour typically observed in light-frame shear walls
(see Figure 7). The few parameters needed to define the
Pivot Cycle, namely the backbone curve and the
variables α 1 , α 2 , β 1 , β 2 for the setting of the unloading
branches, were chosen by calibrating the hysteretic
model on the experimental results of tests carried out at
the University of Trieste in 2005 [15]. Since the cycle is
symmetric, only two parameters were needed: α 1 = α 2 =
4.56 and β 1 = β 2 = 0.2. The yielding force F y was
calculated according to Method A of Eurocode 5 [7], and
the backbone curve was assumed as bi-linear. In Figure 8
it can be seen the cyclic behaviour resulting from the
numerical modelling with Pivot Cycle, compared to the
experimental results.
Figure 9: 3D view of the FE model of the non-isolated
building
The concrete part (ribbed slab and stiff columns) and the
isolating devices were added to the model. The FPS
devices were modelled using a particular non-linear
spring (NLLink) of the SAP2000 library, called “Friction
Isolator”, which was calibrated on the stiffness and
properties of the isolator resulting from the design.
These springs connect the concrete columns to the ribs
of the supporting concrete slab, modelled with elastic
beam elements. The columns are fixed at the base. The
two models are shown in Figure 9 and Figure 10. The 3D
models were used to carry out linear static, modal and
non-linear time-history analyses. The natural vibration
period of the non-isolated building from the modal
analysis was T = 0.67 s.
Figure 8: Comparison between numerical and
experimental cyclic behaviour of a shear wall
5.3 FE MODELLING OF THE BUILDINGS
The two Buildings B and A, with and without base
isolation respectively, were modelled in SAP2000 to
analyze their behaviour under seismic actions. Only the
seismic resistant walls, schematized with the macromodel,
were included in the model. The floor and roof
diaphragms were considered as in-plane rigid. In
Figure 10: 3D view of the FE model of the isolated
building
A time-history analysis was then carried out to compare
the performances of the two buildings. Three
accelerograms generated with the program SIMQKE ver.
4.0 [16], compatible with the response spectrum
calculated according to the Italian regulation for ground

type B and different levels of PGA previously defined,
were used as input for the time-history analyses.
The behaviour of the two buildings was compared in
terms of stresses in the elements and inter-story drifts,
demonstrating the superior performance of the isolated
building. The time history of the interstorey drift in the
isolated and non-isolated buildings resulting from the
dynamic non-linear analysis showed that the maximum
interstorey drift in the isolated building was about eight
times lower than in the non-isolated building. All the
displacement demand is concentrated at the isolation
interface, as can be seen in Figure 11 for the ‘Collapse
limit state”, where it reaches 300 mm. The same figure
demonstrates that the isolator worked (moved) also at
‘Damage Limit State’, providing protection to the
building even for lower intensity earthquake ground
motions.
Figure 11: Displacement in one of the isolating devices
(No. 2) for the CLS and DLS seismic levels
Figure 12 and Figure 13 display the force-displacement
hysteresis loops in a shear wall (see the circled wall in
Figure 1) of the non-isolated building at the 2 nd and
ground floor respectively. The significant interstorey
drift demand causes the plasticization of the shear walls,
especially at the ground floor.
Figure 13: Hysteresis cycles of the shear wall at the
ground floor for ULS seismic action in Building A
On the contrary, all the walls of the isolated building
remain elastic and do not dissipate energy, because the
isolators take the whole input energy.
6 DISCUSSION
By designing separately the isolated (Building B) and
non-isolated structure (Building A) and comparing the
result of the designs, it can be seen that the timber
superstructure of the isolated building can be less strong
and less stiff. The cross-section of the studs in the walls
of the isolated superstructure can be reduced by 33%
(see Table 2). Although this reduction in structural
volume is not dramatic, it still helps cover a part of the
additional cost of the isolation devices. The behaviour of
the two structures, however, is markedly different. In
Building B, the damage is avoided also for strong
seismic events as the timber superstructure is designed
elastically. Since the structural accelerations are very
low, the panic of the occupants and the damage to the
content of the building are prevented. This is a desirable
feature particularly for strategic buildings such as
hospitals, schools, laboratories, etc. where the value of
the content is high, but also for commercial buildings
hosting banks, insurance offices, etc. where the
downtime due to the need of repairing and retrofitting
the building following a strong earthquake ground
motion must be limited as it would lead to significant
economic losses.
Table 2: Comparison between the isolated and nonisolated
structure
Figure 12: Hysteresis cycles of the shear wall at the 2 nd
floor for ULS seismic action in Building A
Cross section of studs
and plates
No. of nails in holddowns
No. of nails in angle
brackets
Nail spacing in panelto-frame
connection
Non-isolated
structure
Isolated
structure
12 cm 16 cm 8 cm 16 cm
45 24
21 15
100 mm 150 mm
Although not explicitly required by current codes of
practice such as the Eurocode, the Italian and New

Zealand regulations, the use of passive base isolation in
residential buildings has notably advantages, both
economical and psychological, as it reduces the
traumatic effects of an earthquake on the occupants, and
avoids injures due to falling pieces and debris.
7 CONCLUSIONS
Timber buildings are widely used around the world.
They are a sustainable construction system and, if well
designed and built, they are suitable for using in seismic
prone areas, because of their low mass and great
robustness. These good qualities are the reason why the
wood structures market is growing. As wood structures
are more and more spread out also in countries with high
seismic hazard, as for example Italy, it is very important
to understand their seismic behaviour and design them so
as the damage is limited if not prevented at all. In this
paper the case study of a 3-storey light-frame timber
building was investigated. Light-frame timber
construction system has the highest dissipative capacity,
due to the many nailed connections between plywood or
other type of sheathing and the light timber frames.
According to the Eurocode 8 and the Italian regulation
for construction, a behaviour factor up to 5 can be used
for this kind of buildings. On the other hand the adoption
of such a high value of q in design may imply significant
damage after the seismic event. This paper investigates
the possibility of using passive base isolation for
preventing this damage.
The 3-storey light-frame case study building was first
designed without and then with seismic base isolation. In
particular, Friction Pendulum system devices were used
for isolating the building. By designing the two
buildings, it was possible to compare the two solutions in
terms of forces, accelerations, damage and costs. The
two buildings were modelled using the commercial finite
element programme SAP2000 and then three seismic
analyses were performed: a linear static, a modal and a
non-linear time history analysis. A research was made to
find the most suitable model for the light-frame shearwalls,
as they are the elements that govern the seismic
behaviour of the entire building. The aim was to find a
simple and reliable model for the structural behaviour.
The model of the wall used for the analyses was also
validated against some experimental results and some
more refined numerical modelling. The parameters for
modelling the shear-walls were calculated based on
prescriptions from Eurocode 5 and New Zealand Timber
Standard.
Both the isolated and the non-isolated buildings were inline
with the seismic requirements of current regulations.
Despite this, during the design earthquake the nonisolated
building undergoes to huge damage, especially
at the ground floor, which is design to dissipate the
seismic energy. Conversely, the isolated structure
remains elastic, as it is designed to remain elastic. The
whole displacement is concentrated in the isolation
devices, which drastically reduce the accelerations in the
superstructure and satisfy the requirements of the
“Damage Avoidance Design”.
As the timber superstructure is very light compared to
the vertical capacity of the isolation devices, the system
is optimized if the timber superstructure is higher than
two storeys and have a smaller plan. For the case study
isolated building, where 16 devices were used, it was
calculated that the cost difference with the non-isolated
building is only 11000 €, corresponding to 0.75% of the
total cost of the non-isolated building. The factors that
influence this difference are three: the saving in wood
material for the isolated structure; the higher cost of the
concrete basement for the isolated structure; and the cost
of the isolating devices. The additional cost of the
isolation is by far counterbalanced by the increased
safety of the building. Therefore, passive base isolation
should be considered more often when designing lightframe
timber buildings in earthquake-prone areas.
However, it should be reminded that the use of passive
base isolation is only possible when the design is
governed by seismic actions and not by wind actions,
which is not always the case. Moreover, the results
obtained in this paper are applicable only for the lightframe
timber construction systems: more research should
be undertaken to cover other types of timber buildings.
ACKNOWLEDGEMENT
The support provided by the Sardinia Region through the
research grant “Numerical-experimental behaviour of
timber panels under in-plane and out-of-plane loading”
is gratefully acknowledged.
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